Keywords

1 Introduction

The sliding zone of a landslide has been recognized as a critical part that controls the initiation, motion, and even morphology of the landslide. Owning to its significance, studies on the properties of sliding zone soil were common throughout the past few decades, with the intention of gaining a better understanding and control of landslides. Traditional testing measures utilized on the sliding zone soil basically consist of shear tests (direct shear and ring shear) and triaxial tests, and most of these tests were conducted as indoor experiments. Indoor experiments have been long recognized due to superiorities such as convenient operation and accuracy control, but flaws inevitably exist, mainly due to the following aspects: a. the samples are structurally disturbed (loss of in situ stress); b. the sample constituents are usually changed by screening to achieve a finer mixture; and c. the sample size. These flaws have raised the concerns for researchers, who have tried various measures to address them. Boldini et al. (2009) conducted large-scale ring shear tests with a DPRI-6 ring shear apparatus developed at the Disaster Prevention Research Institute (DPRI), Kyoto University, to more accurately analyze tsunamigenic landslides. Jiang et al. (2016) emphasized the structure of the sliding zone soil by means of computed tomography (CT) scanning on undisturbed coarse-grained soil and mechanical testing on fine-grained soil. On the other hand, Pirone et al. (2015) and Tang et al. (2014) both intended to reduce experimental errors by combining indoor tests and in situ monitoring to more accurately predict landslides.

Soil deformation develops over time and ultimately ceases, depending on the applied stress magnitude, and this process is time consuming. Consolidation theories of the early ages rarely consider the time factor; however, stress and strain always increase with time. Thus, stress–strain equations that do not consider time cannot accurately describe the soil deformation. In this sense, creep test has become increasingly necessary to investigate soil properties, including deformation traits and constitutive laws. Wahls (1962) conducted sets of soil compression tests under diverse loading rates, and the results showed that the primary consolidation was a time-dependent deformation, revealing that creeping also occurred during primary consolidation. Kabbaj et al. (1988) carried out in situ consolidation tests at various low strain rates and a series of indoor tests at normal strain rates, then discovered that the in situ tests exhibited larger deformations than those of the indoor tests. Notably, the above conclusions were drawn only through creep tests that were conducted with a sufficient time duration. Deformation occurs not only during an increase in stress but also during the consolidation and water drainage under constant stress; this trait of the creep test distinguishes creep from normal deformation. These fundamental discoveries have motivated numerous studies on creep tests in recent years, and deeper knowledge of the sliding zone soil properties and deformation has been achieved (Chang et al. 2021; Liao et al. 2021).

In recent years, studies on soil deformation traits and stress–strain relationships have become increasingly prevalent (Amiri et al. 2016; Miao et al. 2022), as soil behavior under external forces undoubtedly have a significant impact on engineering practices. However, most stress–strain relationships were determined in laboratories, where in most cases, the soil was disturbed, and the sample was remolded. The internal structure of the sliding zone soil mainly originates from the long-time consolidation that makes the soil matrix dense and cohesive; if the soil is remolded, loss of internal structure usually leads to a decrease on the strength. This is noteworthy for the sliding zone soil, which is usually buried deep beneath the slope surface, and therefore the consolidation degree is significant. This question has been discussed by Chen and Liu (2014) who carried out in situ tests and found that the in situ residual strength was less than the peak but greater than the residual strength measured in the laboratory. Similar results were reached by Wen et al. (2016) and Gao et al. (2010), who compared in situ test results with lab test results and reached the same conclusions. Therefore, neglecting the structure destruction would lead to overestimation of the deformation and, as a result, descriptions from the constitutive model may incorporate some errors.

In this paper, the deformation response of sliding zone soil was investigated by means of triaxial testing, in which the soil sample remains undisturbed. This is an innovative approach used to learn more about the soil behavior in its natural environment. To implement this testing scheme, a testing tunnel in the Three Gorges area, China, was utilized as the in situ testing site; at this site, the sliding zone of the Huangtupo 1# landslide is exposed.

2 Materials

2.1 The Huangtupo Landslide

The Huangtupo landslide is the largest reservoir landslide located on the south bank of the Yangtze River in the Three Gorges area of China (see Fig. 1a). The landslide is developed in the strata of the Middle Triassic Badong Formation (T2b2 and T2b3) and is mainly composed of mudstone, pelitic siltstone, and argillaceous limestone. The crown elevation of the landslide is approximately 600 m.a.s.l (meter above sea level), while its toe varies from 50 to 90 m but is submerged by the Yangzte River (Fig. 1c). The landslide has a volume of nearly 70 million m3 and covers an area of 1.35 km2, identified as the largest and most complex landslide in the TGR area of China. The Huangtupo landslide is a large composite deformation body composed of multiple landslides of different stages. It occurs under the influence of specific environmental geological conditions and external dynamic factors and evolves through long-term geological history. Based on previous studies, a landslide geotechnical model is established (Fig. 2), where the effects of rainfall, reservoir water, weak interlayer and cracks on landslide were briefly revealed.

Fig. 1
figure 1

a Location of the study area. b Plan of the Huangtupo landslide. c I-I’ profile of the riverside slump 1#

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Fig. 2
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Geotechnical model and triggering factors

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Four sublandslides can be recognized based on geomorphic features and evolution history in the slope, namely, the Riverside Slump 1#, Riverside Slump 2#, Garden Spot Landslide and Substation Landslide, and the spatial relationships among the sliding surfaces of the four sliding masses were shown in Fig. 1b. Monitoring data (from GPS and borehole inclinometers) indicate that Riverside Slump 1# has deformed the most, meaning that it has the highest risk of failure. Therefore, a group of tunnels under the Riverside Slump 1# were constructed in 2010 (Tang et al. 2015; Hu et al. 2012a, b) to expose the landslide structures, carry out tests, and install monitoring instruments. The tunnel group is composed of a main tunnel and five branch tunnels (BR-1 to BR-5) (Fig. 1b). The test focused in this paper was carried out at the end of BR-3, where the sliding zone soil was exposed (presented in Fig. 1c); other four branch tunnels were not considered for the test mainly due to either not reaching the sliding zone or inconvenient testing environment.

According to the exposure by the main tunnel, the No. l landslide is a typical bedding landslide. The material composition of the landslide mass is mainly gravelly soil and block from the third segment of the Badong Formation (T2b3). Information of the main tunnel reveals that the gravelly soil of the No. l landslide mass is brown red and medium dense. The gravel is argillaceous limestone, light grey, gray yellow and strongly weathered. The particle size of gravel is 5–20 cm with the content about 30–60%. Rock blocks are locally visible, which are composed of argillaceous limestone and limestone of grey-green and blue-grey color. The soil is silty clay, brown red, plastic to hard plastic state, and the ratio of soil to rock is 2:8–3:7. As exposed by the test tunnel and boreholes, the contact between the sliding zone and bedrock is mostly in bedding mode, and the contact interface with underlying bedrock is obvious, where the interface is clear and the phenomenon of cutting layer is locally seen. The boundary between the sliding zone and the overlying landslide body is not clear, showing a gradual transition state. The sliding zone is composed of silty clay with gravel and debris. Its thickness varies from 50 to 100 cm, and the ratio of soil to rock is 6:4 to 8:2. The composition of gravel and fragment is argillaceous limestone with a diameter of 1–5 cm, mostly in the form of grinding circle-sub-angular. The bedrock of the No. l landslide belongs to the third segment of the Badong Formation (T2b3) in the middle Triassic. The rock mass integrity is good, it is medium-thick layer with a thickness of about 20–40 cm. And the landslide and material properties of soil and rocks is shown in Fig. 1c.

2.2 Evolution Mode of the Landslide

Figure 2 presents a schematic of evolution process of the Huangtupo landslide (Tang et al. 2015; Deng et al. 2017). Along the axes of the Guandukou Syncline (Fig. 3a), the dip-slope of the Huangtupo landslide emerges and evolves into a gravitationally unstable state. The slope will keep stability to avoid fully sliding until it is deformed by the precedent long-term gravitational deformation (PLTGD) of toppling and deep-seated creep (Fig. 3b). The PLTGD gradually reduces the integrity and strength of the rock mass, and this reduction process will continue and deteriorate under the role of the river cut. When the integrity and strength of the rock mass are reduced to a certain extent that insufficiently maintains the slope stability, large-scale landslides occur, which we call pre-landsides as shown in Fig. 3c. Subsequently, the Huangtupo landslide undergoes long-term dismantling under the influence of river erosion, rainfall, and river water fluctuation (Fig. 3d).

Fig. 3
figure 3

Evolution mode of the Huangtupo landslide

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3 Methodology

3.1 Instrument Employed for the Test

The YXSW-12 instrument was used for the test, which was developed jointly by the Changjiang River Scientific Research Institute (China) and Zhaoyang Experimental Instruments Co., Ltd. (China). The testing system comprises. a. components such as a vertical loading unit (jack), horizontal loading unit (hydraulic pressure pillow), deformation gauge and counterforce frame, as plotted in Fig. 4, and b. a power and control unit, composed of a personal computer (PC) that acts as a command center and collects data, an EDC-series servo driver for converting and transmitting the signals, and a servo valve, booster and an oil source for supercharging the vertical and horizontal loading units.

Fig. 4
figure 4

Structural representation of the YXSW-12 instrument

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Specifically, the hydraulic pressure pillow expands when charged with oil and exerts forces on the soil sample. Horizontal deformation could occur as the pillows would be compressed to deformation under the forces from the creeping soil; during this process, the horizontal pressure would be held constant by the computer-servo system. Specifically, the applied stress would be kept constant using a system that has a stress compensation system. The system runs with the connected line of PC, the EDC, the servo valve and the booster. When the actual stress changes because of external factors, e.g. soil compression and horizontal expansion, the compensation system will automatically detect and then compensate or discharge the pressure by adding or discharging the used oil. The system has also been successfully employed in previous tests by its creation team (Zhang et al. 2011).

The tested cuboid sample had a height of 1000 mm and a cross-sectional area of 500 mm × 500 mm, as presented in Fig. 3. The soil was cut out of the horizontally enveloping soil blocks using a procedure that reduced the disturbance to the soil, and the bottom surface of the cuboid remained connected to the base soil. The four vertical planes of the sample were oriented at 35°, 125°, 215° and 305°, wherein the 35° plane was approximately oriented parallel to the moving direction of the landslide (as plotted in Fig. 5).

Fig. 5
figure 5

Orientations of each testing surface and the landslide

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3.2 Loading Scheme and Test Result

The loading scheme is described using a flow chart, presented Fig. 4. For the vertical stress series, it was set according to the overlying load of the sliding zone soil; for the lateral stress, it was determined by in situ measurement. The confining pressure was set according to the measured initial ground stress. Therefore, the confining pressure was kept at 0.5 MPa during the test, and the vertical pressures of 0.65 MPa, 0.80 MPa, 0.95 MPa and 1.10 MPa were sequentially applied. The loads, starting with stress level 1, were instantaneously applied on the top surface of the sample, and the subsequent level of stress would be applied when the increase in displacement approximately ceased.

4 Results

4.1 Deformation Results and Characters

Deformations of the five sample faces were simultaneously detected by deformation gauges (marked in Fig. 4), and the resultant curves are plotted in Figs. 6, 7 and 8. Figure 6 presents the displacement development under stress level 1, in which each face first undergoes an instantaneous elastic deformation and then undergoes an attenuating creep.

Fig. 6
figure 6

Displacement–time curves of the 5 surfaces under stress level 1. Negative values indicate shrinking deformation, which is discussed in the section “Anisotropic deformation”

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Fig. 7
figure 7

Displacement–time curves of the 5 surfaces under stress level 2

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Fig. 8
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Displacement–time curves of the 5 surfaces under stress levels 3 and 4

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Specifically, the 35° face quickly deforms by 7 mm in approximately 15 h and then begins an attenuating phase until reaching a steady value at 17.3 mm. The 125°, 215° and 305° faces deform much like the 35° face, although the magnitudes vary; the total displacements are 1.2 mm, 5.9 mm and 6.3 mm, respectively. Reasons giving rise to the anisotropic deformation traits of the surfaces would be later discussed in the section “Anisotropic deformation”.

The vertical deformation curve from this test also undergoes a linear increase followed by an attenuating creep phase, but the linear phase continues for a much larger time than that of the other faces, nearly 350 h. In total, the vertical displacement during stage 1 (under stress level 1) is 13.6 mm. This phenomenon is unique and can be explained by the following observations. The unsaturated soil undergoes water drainage and volume change under applied forces; therefore, the stress would be transferred to the soil matrix as the water drains out, and the deformation ultimately ceases. This represents a general consolidation process. In this test, the transient vertical loading on the sample induces a fast pore pressure increase, which concentrates near the loading surface.

As shown in Fig. 4, stress levels 2 and 3 were applied vertically and in succession when the displacement in the former stage had approximately ceased. Figure 7 presents the displacements of all the faces in stage 2 (stress level 2), and Fig. 8 shows the results from stages 3 and 4 (stress levels 3 and 4). These curves are plotted independently for clarity, as the change in displacement during stages 2–4 are very small. The greatest increases in the deformation curves for stages 2–4 are only approximately 1 mm. Notably, stress levels 3 and 4 respectively equals and exceeds the initial in-situ stress calculated at the embedment depth of the sliding zone. Therefore, the deformation curves generally exhibit attenuation (Wen and Jiang 2017), indicating that only creep happens, and soil failure will not occur no matter how long the creep continues.

4.2 Analysis on the Low Compressibility of the Sample

Two factors mainly account for the low compressibility: a high preexisting compression degree of the soil and a high rock content.

  1. (a)

    High preexisting compression degree

For clay samples, the primary consolidation that normally occurs due to external pressure would stop once the excess pore water pressure has dissipated. However, continuous forces on the soil would ultimately lead to secondary consolidation, enhancing interparticle forces and solidifying some cementing constituents. Clay that undergoes secondary consolidation usually exhibits a higher compression modulus and bears capacity in engineering applications, behaving like “old clay,” which refers to clay that formed earlier than the late Pleistocene (Li 2004).

Figure 9 presents two void ratio (e) reduction curves (blue) developing with various geological times for the same clay. In Fig. 9, P0 is the applied force for normal consolidation, and e0 is the final void ratio; Pcq is a greater stress applied on identical soil, and e0′ equals e0. The results indicate that compression to T0 under P0 takes 10 thousand years; however, under the higher pressure Pcq, the time required to reach the same e value (Tcq) decreases to less than 0.1 year. Similarly, Pcq and P0 curves can serve as a laboratory normally consolidated soil case and a historical consolidation case, respectively. A greater force is needed in the laboratory to reach the density equal to that in historically consolidated soil. We call such historically consolidated soil quasi-over consolidation (QOC) soil, with its quantitative parameter defined as:

$$ QOCR = P_{0} /P_{{{\text{cq}}}} $$
(1)
Fig. 9
figure 9

Void ratio (e) reduction curves developing with various geological times, where NCL denotes the natural consolidation line (based on Li 2004)

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where QOCR denotes the quasi-over consolidation ratio.

The Riverside Slump 1# has a deep-seated sliding zone; consequently, the compressibility of the sliding zone soil tends to be smaller than that of a normally consolidated soil from a laboratory experiment, as QOC can create a denser soil structure.

  1. (b)

    High rock content

Rock content considerably affects soil compressibility. Xu et al. (2009) discovered, after testing samples with rock contents of 30%, 40%, 50% and 60%, that the uniaxial compressive strength (UCS) of soil samples increases with the rock content. Tests by Li and Wang (2004) showed that gravelly soil with low rock contents (<30%) behaved like a fine-grained soil; conversely, gravelly soil with high rock contents (>60%) behaved like a rock mass.

In this test, the rock blocks generally have a sparse to dense distribution from the sample top to the bottom, and the total rock content (particle diameter > 5 mm) reaches 54.3% in the sample cube. Therefore, the cube is characterized as gravelly soil with inhomogeneously embedded rock blocks. As demonstrated by previous studies, the compressibility of soils decreases with the increase in rock content in the soil mass, due to the following reasons: a. When neighboring rocks obstruct each other during shearing, the soil movement is locked. b. Peculiarly located rocks have the potential to block the development of shear zones and cracks. Some test results suggest that once the rock content exceeds 50%, the largest rocks in the soil generally determine the shear zone shape and location, and mechanical properties (Xu et al. 2009).

4.3 Constitutive Law

The semi-empirical and semi-theoretical model, which combines element model and empirical model, is established based on the analysis of the creep characteristics of sliding zone soil samples in the in situ test, and can describe the nonlinear creep characteristics of sliding zone soil well. As indicated by Figs. 6, 7 and 8, the in-situ compression creep test exhibits typical attenuated creep characters, where total strain is normally separated into the linear viscoplastic element and the nonlinear viscoplastic element. Specifically, the linear viscoplastic element is reflected by the assembly of rheology bodies frequently used for studying soil behavior; empirical equations will be introduced to describe the nonlinear viscoplasticity, which cannot be described by the rheology bodies.

Viscoelastic deformation is not incorporated in the constitutive equation as an independent term because it is not clearly displayed in the deformation data, which usually exhibits apparent linearity. Reasons of lacking the viscoelastic deformation may well the due to the pre-existing vertical pressure on the sample from the overlying soil, which had lasted long during the historical period. Elastic deformation already occurred in the historical period, but no obvious rebound took place owning to the creep generated by the long-time vertical load acting on the soil. Thus, the long-time and high-intensity compression on the soil caused irreversible creep that may account for a considerable proportion of the deformation. Consequently, the configuration is shown in Fig. 10, where a term of linear viscoplasticity and a term of nonlinear viscoplasticity are connected in series (Wang 2007; Zheng 2016); notably, the linear viscoelasticity term (dashed) is excluded from the constitutive equation.

Fig. 10
figure 10

Configuration of the constitutive equation. The term inside the dashed rectangle is eliminated

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4.4 Expression of the Constitutive Model

The constitutive equation is deduced based on the element configuration given in Fig. 10. Based on previous studies, the linear viscoplasticity term can be written as:

$$ \varepsilon_{l} = \frac{{\sigma - \sigma_{s} }}{{E_{l} }}\left( {1 - \exp \left( { - \frac{{E_{l} }}{{\eta_{l} }}t} \right)} \right) $$
(2)

where \(E_{l}\) and \(\eta_{l}\) denote the elasticity modulus and viscosity coefficient of the rheology bodies, respectively. \(\sigma_{s}\) denotes the yield stress of the soil.

On the other hand, the nonlinear viscoplasticity term can be expressed by an empirical model that reflects such deformation traits. Empirical models frequently adopted for describing the rheological behavior generally include the power function, exponential function and logarithmic function. Specifically, the power function is usually used for reflecting the decaying creep; the exponential function, for steady creep; and the logarithmic function, for accelerating creep. In this sense, the power function is selected to describe the decaying creep in this test, expressed as

$$ \varepsilon_{n} = \left( {\frac{{\sigma - \sigma_{s} }}{A}} \right)^{m} $$
(3)

where \(m = 1/m^{\prime}\) and \(m > 1\).

Nonlinear viscoplastic creep relaxes with time; thus, \(A\) and \(m\) in Eq. (3) should be time-related parameters, which can be written as:

$$ A = A(t),\quad m = m(t) $$
(4)

\(A(t)\) can also be represented by a power function (Sun 1999), which can be expressed as:

$$ A(t) = A_{0} t^{ - \alpha } $$
(5)

Then, Eqs. (3) and (5) are combined so that the nonlinear viscoplasticity term can be finally obtained:

$$ \varepsilon_{n} = \left( {\frac{{\sigma - \sigma_{s} }}{{A_{0} t^{ - \alpha } }}} \right)^{m} = \left( {\frac{{\sigma - \sigma_{s} }}{{A_{0}^{{}} }}} \right)^{m} t^{\beta } $$
(6)

where \(\beta\) is the creep index, \(\beta = m\alpha\), and \(A_{0}\) is the nonlinear coefficient of the deformation.

Therefore, the final expression of constitutive law can be obtained through series connection of two terms (Wang 2007; Zheng 2016), with the final relationship obtained as

$$ \varepsilon = \frac{{\sigma - \sigma_{s} }}{{E_{l} }}\left( {1 - \exp \left( { - \frac{{E_{l} }}{{\eta_{{_{l} }} }}t} \right)} \right) + \left( {\frac{{\sigma - \sigma_{s} }}{{A_{0}^{{}} }}} \right)^{m} t^{\beta } $$
(7)

As indicated by Eq. (7), all parameters \(\sigma\), \(\sigma_{s}\), \(E_{l}\), \(\eta_{l}\), \(A_{0}\), m and β are included. The \(\sigma\) (deviatoric stress) was calculated to be 0.15 MPa by vertical pressure subtracting confining pressure. \(\sigma_{s}\) is evaluated by fitting the compression model with test results, as this parameter is decided by viscoelastic deformation of testing curve, which is considered nonexistent for the tested soil. \(E_{l}\) (elasticity modulus) and \(\eta_{l}\) (viscosity coefficient) were recognized from the initial linear segment of the vertical deformation curve. The remaining parameters, namely \(A_{0}\), m and β, were obtained by fitting with the Origin and the result is shown in Fig. 11. The coefficient of determination (r2) turns out 0.9768, indicating that the created model is convictive to match the test data. Therefore, Eq. (7) is ultimately determined to interpret the constitutive law of the tested sample.

Fig. 11
figure 11

Fitting result for the intact data during stress level 1

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It is noteworthy that Eq. (7) is verified employing only the data of the first stage (stress level 1), mainly because those data possess manifest linear and nonlinear courses of deformation, and thus the random errors can be shrunk to the minimum. Further, Eq. (7) is utilized to fit the data respectively of stages 2, 3 and 4, and the data prove to be ideally fitted as well (not displayed in the text).

5 Discussion

5.1 Anisotropic Deformation

The displacement–time curves presented in Figs. 6, 7 and 8 indicate that strain is directionally favoring the 35°, 215°, 305° and vertical surfaces, respectively. Specifically, the 35° surface expands the most among the horizontally oriented surfaces, expanding approximately 3 times the amount of displacement at the 215° and 305° surfaces.

When compressed by vertical stress, the high rock content in the soil would enable the sample to behave like a rock mass. That is, cracks may emerge at locations that have concentrated stress and at the soil-rock interfaces, where the cementation is weakest. For the tested sample, the areas of concentrated stress are generally oriented along the \(45^{ \circ } + \phi /2\) direction (marked in Fig. 12), along which the generated shear strength is generally the greatest for a compressed cube. Additionally, as the rock content (particle diameter > 5 mm) reaches 54.3% of the soil-rock mixture, the numerous soil-rock interfaces in the sample induce cracks widely distributed throughout the sample.

Fig. 12
figure 12

Schematic diagram showing the deformed 35°-oriented surface of the cube and structures within the cube after testing

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The 35° surface is similar to the landslide orientation (see Fig. 5). As has been previously discussed, particles within the sliding zone would gradually align with the sliding direction after long-time creep and movement of the soils and rocks. Therefore, cracks that form at the soil-rock interfaces would also be oriented at 35°, possibly inducing deformation towards this orientation. This explains why the 35° surface deforms much more intensely than the other surfaces.

The 215° surface, however, presents a shrinkage trait (negative displacement in Figs. 6, 7 and 8) under the joint action of a vertical load and constant confining pressure of 0.5 MPa. With respect to the cracks mentioned above, because the soils and rocks generally move towards the orientation of 35°, a shrinkage in that orientation is expected. The 125° and 305° surfaces are also expected to be characterized with a distribution of cracks, which would result in an inhomogeneous shear strength under compression; specifically, orientations of the main cracks (35°) may have components to the orientation of 125°, which is estimated from the intersection angle between the sliding direction and the 35° orientation (plotted in Fig. 5). Therefore, the deformations of these surfaces would be very different.

5.2 Landslide Movement Related to the Sample

The Riverside Slump 1# has been monitored since 2003; various measures including the use of GPS, borehole drilling, and time-domain reflectometry (TDR) have been employed in this monitoring effort. Some monitoring sites on the Riverside Slump 1# are presented in Fig. 1b, c; these sites are distributed not far from tested sample. The monitoring measures shown in Fig. 1b consist of GPS sites and boreholes; specifically, the GPS sites are responsible for detecting deformation of the slope surface, and boreholes were drilled into the sliding zone and instrumented to detect the deformation of the sliding zone. Monitoring data are presented in Fig. 13.

Fig. 13
figure 13

Time series monitoring data from 2003 to 2011

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G9 is located above the sample, and the average deformation velocity at this location is 22.97 mm/year, while the average deformation velocities for the data curves in Fig. 13 range from 13.71 mm to 26.78 mm/year. HZK5 extends to the sliding zone soil positioned very close to the tested sample, as clearly marked in Fig. 13 with a dotted line. The calculated deformation velocity at HZK5 is 10.65 mm/year.

The displacement developed under stress level 3 (see Fig. 8) is selected to represent the deformation velocity of the soil cube because the stress applied in the third stage (0.95 MPa) equals the in situ overlying load, which can be estimated from the burial depth and the specific gravity of the tested sliding mass. At this time, the sample enters into a sustained nonlinear viscoplastic phase, consistent with the ordinary state for soil creeping. The calculated velocity is 5.84 mm/year for the 35°-oriented surface, whose normal direction is basically parallel to the landslide direction, as explained in the previous section.

The deformation laws of the soil cube and the sliding zone are compared, as a similar stress environment was considered for both the confining pressures and the vertical loads. Nonetheless, a distinction of approximately 5 mm appears between the detected velocity from HZK5 and the calculated velocity for the sample, even though they are located very close to one another. This phenomenon may be due to external factors, namely, rainfall and water level fluctuations. The impacts of these factors have been demonstrated to be significant for the Huangtupo 1# landslide. Wang et al. (2016) investigated the roles on the deformation of the Huangtupo 1# landslide played by rainfall and water level fluctuation, with the results indicating that approximately half of the total deformation was attributed to rainfall and water level decrease. Therefore, the final velocity to be evaluated must take rainfall and water fluctuation (presented in Fig. 13) into account.

The data indicates, to some extent, that deformation decreases from the landslide surface to the sliding zone. In addition, the landslide undergoes creeping without a high risk of abrupt failure, as determined by the attributes of the sliding zone soil. As previously analyzed, the sliding zone possesses very low compressibility due to its high rock content and high degree of historical consolidation. The applied vertical force, which is even higher than the original stress induced by the overlying soil, cannot cause the abrupt failure of the sliding zone soil. That is, the Huangtupo 1# landslide has been creeping for years, and this will not change under normal circumstances. Nonetheless, extreme conditions for rainfall and water fluctuation in the Three Gorges Reservoir area must be considered, as they may generate problems such as a considerable increase in the hydrodynamic pressure and water content, possibly leading to failure.

However, further discussion is still requested on the topic, that since the sample tested is situated near the mid location of the sliding surface, the corresponding traits may be not representative enough for the intact sliding zone. On this issue, some discussions would be further given in the followings:

  1. (a)

    The intact sliding zone is embedded beneath the same material of the landslide mass, which can be clearly seen from Fig. 1c. In this sense, it is reasonable to consider that the whole sliding zone soil has no much difference in terms of the deforming traits and mechanism compared with the tested sample.

  2. (b)

    The landslide has been creeping for years according to data from the monitoring sites, which are installed widely on the landslide surface (Wang et al. 2016). Therefore, creeping keeps occurring throughout the landslide, which is a general character for the whole sliding mass and zone. In this sense, behavior of other locations can be positively analyzed based on the results of this study.

6 Conclusions

An in situ triaxial creep test was carried out for exploring the properties of the sliding zone soil, which is characterized by a high rock content and high degree of compression. The tested sample was located in the tunnel that passes through Riverside Slump 1#, a subdivision of the Huangtupo landslide. The test was completed over 191 days, and the results were then analyzed. The conclusions are as follows:

  1. (a)

    The tested sample exhibits very low compressibility due to the high rock content (54.3%) and long-term compression exerted from the overlying soil. The dense structure leads to a high strength and stability, resulting only the decaying creep of the creep test.

  2. (b)

    The constitutive model of the vertical stress–strain-time relationship is discussed and deduced. Linear viscoplasticity and nonlinear viscoplasticity terms are used in series to establish the equation, with which an empirical model (power function) is combined to more accurately describe the nonlinear viscoplasticity. The results show a strong fitting relation between the testing data and the constitutive model.

  3. (c)

    The anisotropic traits of the horizontal deformation of the sample are discussed. The surface at an orientation of 35° generates much larger displacement than those at the other surfaces, possibly attributable to the alignment of cracks within the sample cube.

The monitored data from HZK5 behave similarly to the measured data from the 35° surface in terms of the velocity of displacement. The data from the GPS sites and boreholes indicate that the Riverside Slump 1# has been creeping over the past years. This creep may be controlled by the properties of the sliding zone: decaying creep occurs and no abrupt failure occurs in the tested sample when stress magnitudes similar to those of the in situ stress environment are applied. However, unavoidable limitations of the test and extreme external factors, such as unexpected rainfall and water fluctuation in the Three Gorges area, cannot be neglected when evaluating the long-time stability of the Huangtupo Riverside Slump 1# landslide.